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Investigation of the Collapse of the Chirajara Bridge A summary of a study conducted by members of ACI Committee 133, Disaster Reconnaissance by Santiago Pujol, Michael E. Kreger, Jonathan D. Monical, and Arturo E. Schultz

T

he government of Colombia is conducting an expansion of its built infrastructure, including the transportation system, through its Agencia Nacional de Infraestructura (National Infrastructure Agency, ANI). The Chirajara Bridge was one of 47 bridges in a project1 to expand the highway from Bogotá to Villavicencio from two to four lanes, with the bridge carrying two of those lanes. While under construction, one-half of the Chirajara Bridge collapsed on January 15, 2018. The mission of ACI Committee 133, Disaster Reconnaissance, is to collect information from disasters affecting reinforced and prestressed concrete structures. The information collected is used to inform ACI committees that develop code provisions, design recommendations, and specifications. Following the collapse, a representative of ACI Committee 133 contacted ANI to request access to the site to collect any information that could link the collapse to needed improvements in design recommendations. This article summarizes the findings of the investigation.

The exterior spans were approximately 80 m (262 ft) long. At the terminal ends of these shorter spans, massive abutments were placed to provide reactions against a portion of the loads imposed on the longer center span. The West Tower (at Axis B in Fig. 1) collapsed, claiming A

B

80.00

C

286.30

D

(a) A

B

80.00

The Bridge Overview

The Chirajara Bridge was a cable-stayed bridge, with a total projected length of 446.3 m (1464 ft), spanning the nearly 150 m (500 ft) deep Chirajara gorge. It was supported by two reinforced concrete towers, each to have 52 stay cables. Each tower was 107.34 m (301 ft) tall (above a drilled shaft foundation), and the center-to-center distance between the towers was 286.3 m (938 ft). The stay cables consisted of bundles of ASTM A416/A416M Grade 270 (270,000 psi [1860 MPa] tensile strength) strands. The typical spacing of stay anchorages along the length of the bridge deck was 9.25 m (30 ft), shown in Fig. 1.

80.00

Stays not in place

7

(b)

6

5

4

3

2 1

1

2

3

4

5 6

7

8

9 10 11 12 13 14 15

129.5 m completed deck

Fig. 1: The Chirajara Bridge was a cable-stayed bridge, with a total projected length of 446.3 m (1464 ft), spanning the nearly 150 m (500 ft) deep Chirajara gorge: (a) schematic of the completed bridge (viewed from the south); and (b) schematic of the West Tower at the time of the collapse on January 15, 2018 (Note: Dimensions are in m; 1 m = 3.3 ft) www.concreteinternational.com | Ci | JUNE 2019

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the lives of nine workers. At the time, bridge construction was near completion, with the ends of the deck that cantilevered from each tower being approximately 30 m (98 ft) from one another. No construction activities involving heavy weights or large forces are reported to have been taking place on the collapsed portion at the time of failure.

Key components, dimensions, and properties

The deck was a 13 m (43 ft) wide, composite steel-concrete structure with two 1500 mm (59 in.) deep steel longitudinal girders and 840 mm (33 in.) deep steel transverse beams supporting a 200 mm (8 in.) thick reinforced concrete slab every 3 m (9.8 ft). Each tower had a hollow 38 m (125 ft) tall mast to anchor the stay cables (Fig. 2). The mast had a rectangular section and was supported by a structure with a diamond-shape elevation; the widest part of the diamond accommodated the roadway. The specified compressive strength for the tower concrete was 35 MPa (5000 psi). Standard cylinder test reports provided to the supervision company show results exceeding the specified strength. The outer dimensions of the cross section of the mast were 6 m (20 ft) in the longitudinal direction of the deck and 4.6 m (15 ft) in the transverse direction. The corresponding inner dimensions were 4 and 3.8 m (13 and 12.5 ft), respectively. The walls were therefore 1000 and 400 mm (39 and 16 in.) thick.

0.60

4.60

4.60

0.60

1.00

1.00

6.00

1.00

Columns supporting the mast were 6 by 1.6 m (20 by 5.2 ft) in cross section (solid), with the longer side oriented in the longitudinal direction of the deck. The lower part of the tower included a 500 mm (20 in.) thick wall acting as a web between the columns (Fig. 3). At the top of this web and below the deck was a 6 m wide, 600 mm (24 in.) thick slab (Fig. 4). The slab acted as a tie between the columns at the elevation with the widest separation between the columns. This element was referred to as the “tower slab” in the drawings. The standard to be met by the deformed reinforcing bars was not specified in the design drawings provided to the authors. Nevertheless, mill certificates provided to a supervision company for the highway expansion project listed minimum and maximum yield strengths of 60,000 and 78,000 psi (420 and 540 MPa), respectively, with a required minimum elongation of 14%. This indicates that the bars were specified as ASTM A706/A706M Grade 60. According to the drawings, longitudinal reinforcement in each column comprised 146 No. 10 bars (the drawings indicated the bar size using ASTM A706/A706M nomenclature), resulting in a reinforcement ratio of 1.2%. The volumetric transverse reinforcement ratio was 0.5%. Longitudinal (vertical) reinforcement in the mast comprised 156 No. 10 bars near the outer perimeter face of the hollow section and 46 No. 6 bars near the inner perimeter face, resulting in a reinforcement ratio of 1.1%. The volumetric ratio of transverse reinforcement in the mast also was 1.1%. The interior of the mast was lined with steel plates. The plates 0.075

15 No. 10 at 0.10

0.075

38.0

0.40

No. 3 crossties No. 4 tie at 0.20 70.00

1.40

2.15

No. 6 bar at 0.20 No. 8 vertical bar at 0.20 (72 total each face of web)

14.15 17.62

No. 4 bar at 0.2

8.50

8.50

Fig. 2: Schematic views of a bridge tower (Note: Dimensions are in m; 1 m = 3.3 ft)

30

0.050

37.34

32.34 5.00

4.00 0.25 2.38

0.5

1.60

Tower slab Web

58 No. 10 at 0.10

13.00

11 No. 4 crossties at 0.20

6

32.00

1.60

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1.6

Fig. 3: Section through tower column and web (Note: Dimensions are in m; 1 m = 3.3 ft; bar sizes are indicated using ASTM A706/A706M nomenclature— roughly, diameters in multiples of 1/8 in.)

14 tendons of eight (8) 0.60 in. diameter strands

No. 4 at 0.25 L = 2.95

Column

No. 4 at 0.25 L = 2.95 0.60

0.10

No. 4 at 0.25 L = 12.00

12 tendons of one (1) 0.60 in. diameter strand

57 No. 4 at 0.25 L = 5.90 both faces

3 No. 4 at 0.30 L = 1.75

No. 4 at 0.25 L = 12.00

Column

Web

Fig. 4: Section through slab that acted as a tie between the columns at the elevation with the widest separation (Note: Dimensions are in m unless noted otherwise; 1 m = 3.3 ft; 1 in. = 25 mm)

were 20 mm (0.8 in.) thick in the direction of the deck and 25 mm (1 in.) thick in the perpendicular direction, and they were anchored to the concrete with headed studs. The web connecting the lower columns of the tower was reinforced with vertical and horizontal bars near each face. The vertical reinforcement comprised No. 8 deformed bars (1% reinforcement ratio) and the horizontal reinforcement comprised No. 4 bars (0.26% reinforcement ratio). The drawings showed No. 4 crossties at approximately every third

vertical bar (as shown in Fig. 3). While the drawings also called for a spacing of 200 mm (8 in.) for both the vertical and horizontal bars, we observed in the remains of the West Tower web that the vertical spacing of the No. 4 bars varied between 100 and 200 mm (4 and 8 in.). A design section shows that the horizontal web reinforcement was terminated at hooks anchored at exterior longitudinal bars in the column (Fig. 3). The tower slab reinforcement is shown in Fig. 4. Reinforcement anchored in the columns consisted of twelve

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0.6 in. (15 mm) diameter unbonded post-tensioning strands. We believe that these were specified as ASTM A416/ A416M Grade 270 strands. Notice that the deformed reinforcement was terminated near each column face. Because the construction process resulted in cold joints between the concrete in the tower slab and the columns, the deformed bars would have been incapable of transferring tension forces between the tower slab and the columns. Deformed reinforcing bars were lap spliced. Splices were located away from critical sections.

Construction sequence

After the towers and abutments were constructed, the deck was built from each abutment to the adjacent tower. Shoring supported the deck during this operation. Deck segments were added from each tower toward the center of the middle span. These segments were supported by stay cables, and the weight of each was balanced by a corresponding segment on the opposite side of the tower supporting it. Once the initial seven pairs of stay cables were installed for each tower, additional segments in the center span were balanced by the stays anchored in the nearest abutment.

Collapse

Reported conditions

For the portion of the bridge that collapsed (supported by the West Tower, on Axis B), 13 pairs of stay cables are reported to have been installed from the tower toward mid-span, with the total length of deck cantilevered approximately 129.5 m (425 ft) from the center of the tower (shown in Fig. 1). We understand that the 200 mm thick reinforced concrete deck slab had been cast along this entire length. We also understand that the deck supported by the collapsed tower contained only small live loads at the time of collapse. On the deck carried by the (remaining) East Tower, 12 pairs of stay cables had been installed, and deck framing had been installed for the 13th deck segment. Nevertheless, no

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Fig. 5: Frames from a video of the collapse of the West Tower, recorded by a nearby security camera north of the bridge. The time interval between each frame is approximately 1 second

concrete had been placed for the 12th and 13th deck segments. It is also important to review what the structure was not experiencing at the time of collapse. Neither portion of bridge was carrying significant live load, and the concrete deck did not yet have a planned asphalt wearing course. Lastly, there were no seismic demands, no reports of strong winds, and no indication of foundation distress at the base of the towers.

Video evidence

A security camera recorded the West Tower during the collapse. Frames extracted from this video are shown in Fig. 5. The sequence shows clearly that the dominant feature of the collapse involved separation between the columns on opposing sides of the tower, suggesting a tensile failure at or near the tower slab. The other bridge components fall almost directly downward.

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Field Assessment The debris

The displacements seen in the video simplified the task of identifying and labeling the structural members in the debris field (Fig. 6). Examination of the debris revealed that failure involved separation between the lower columns and web. Figure 7 shows that the horizontal No. 4 bars in the web fractured at the juncture with a column. Despite the collapse, the remains of the web had limited cracking, indicating that strains were concentrated near the columns, where the horizontal web reinforcement fractured. We observed no failures in stay-cable anchorages in the bridge deck, and we noted that deck girders fractured at splice locations.

The remaining tower

On January 25-26, 2018, the East Tower was still in place as constructed. We inspected the tower with a powerful

scope, cameras with zoom lenses, and an aerial robot (drone) (Fig. 8(a)), and we observed that the cold joints at the ends of the tie (or tower slab) were open (Fig. 8(b)). We also noted a large crack in the north side of the web, starting near the cold joint and extending toward the foundation. We estimated the crack to be 8 to 9 m (26 to 30 ft) in length (Fig. 8(c)).

Analyses of the Towers

Simple idealization of the tower as a truss

The video of the collapse suggests that the failure was related to gravity demands. We made the following estimate of the vertical load acting on the West Tower at the time of the collapse. Figure 9 shows a free-body diagram of the West Tower and deck at the time of the collapse. The distributed deck load was about 6.5 kPa (135 psf), including the concrete slab and stay cables. The deck cantilevered 129.5 m (425 ft) from the West Tower, so this deck portion totaled about 129.5 m × 13 m × 6.5 kPa = 11,000 kN (2.5 million lbf). Because the 80 m (262 ft) span measured to the outermost stay of the exterior span was shorter than the cantilevered portion, the group of stay cables anchored in the abutment must have been stressed to balance the system. The vertical component of forces in these stays would have added to the vertical forces acting on the

Fig. 6: View of collapse site captured using an aerial robot (drone). LCS and LCN indicate lower column south and north sides, respectively; and UCS and UCN indicate upper column south and north sides, respectively (images courtesy of Xenital S.A.S.)

(a)

(b)

mast. We were not provided with records of the cable-stressing operations. However, we estimated this component to be 3600 kN (820,000 lbf) by assuming that the horizontal component and the concrete transition block (discussed in the next paragraph) balanced the cantilever load. The deck in the exterior span totaled about 66 m × 13 m × 6.5 kPa = 5600 kN (1.26 million lbf), where the 66 m (217 ft) dimension is the approximate length of the deck from the tower center toward the abutment. A concrete transition block (about 3 x 3 m [9.8 x 9.8 ft] in cross section and 15.7 m [51.5 ft] in length) had been placed between the deck and the abutment. This block included voids and weighed about 2220 kN (500,000 lbf), and half of the weight was supported by the end of the deck. The total force delivered by the stay cables to the mast atop the West Tower was therefore 11,000 + 3600 + 5600 + 2220/2 = 21,300 kN (4.79 million lbf). The mast itself weighed nearly 12,000 kN (2.7 million lbf). The total vertical force introduced by the mast to the upper columns was therefore 21,300 + 12,000 ≈ 33,300 kN (about 7.49 million lbf). Lastly, the columns in the upper portion of the tower weighed nearly 8000 kN (1.8 million lbf) each. If one idealizes the upper half of the tower as a simple arch (that is, a two-element truss), with the weight of its elements being applied in halves at element ends, the horizontal component of the axial force in each of the upper columns would be approximately

(c)

Fig. 7: Views of the lower column south (LCS), showing the remnants of horizontal No. 4 bars at what had been the junction with the web: (a) overall view; (b) view of bars that were bent downward as the column fell away from the web; and (c) closeup of bars that necked down and fractured at the juncture www.concreteinternational.com | Ci | JUNE 2019

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  8000 kN   1 1   × × = 4130 kN ≈ 928, 000 lbf 33,300 kN + 2  2   2 5 

The factor 1/2 is the inverse of the number of columns in the tower. The factor 1/5 is approximately the tangent of the angle between the column and the vertical. The weight of 8000 kN is multiplied by 2 because there were two columns, and it is divided by 2 to follow the common and stated assumption that the weight of a truss element can be assigned in halves to the nodes at its ends. The estimated force had to be resisted mostly by the tower slab that acted as a horizontal tie between the columns. The deck itself did not help resist this tension because the only direct connection with the tower occurred through bearing pads meant to support at least a fraction of the deck segment directly over the tower. In the estimates given previously, it was assumed that this fraction was negligible relative to the other weights involved.

(a)

(b)

(c)

Resisting a critical tensile force of nearly 1 million lbf (4450 kN)—designing mechanisms to transfer it from one element to another, ensuring reliable long-term capacity, and prohibiting cracking in concrete—would require extreme care in the design and meticulous detailing of the reinforcement. Idealizing the entire tower as a truss (that is, ignoring the web) would nearly double the estimated axial force in the slab that acted as a tie near tower mid-height (to almost 2 million lbf [8900 kN]). In contrast, the only continuous reinforcement in this element comprised twelve 0.6 in. diameter, Grade 270 unbonded strands. The nominal strength of these strands would have been about 12 × 0.217 in.2 × 270 ksi = 0.7 million lbf (3100 kN), which is much less than the 1 to 2 million lbf tensile force needed to hold the two columns in position at the widest point in the tower. Therein lies the most glaring issue that may explain the collapse. The tower slab—the element that acted as a tie at mid-height of the tower—did not have sufficient reinforcement. The load on the bridge was smaller than what was expected at service. There was no reason for any stresses in the system to be remotely close to limiting values. Yet, statics and the simple bridge representation described previously show that the mid-height tower slab acting as a tie to resist a force did not have adequate capacity. From that point of view, the reason for the collapse should not be difficult to understand.

Fig. 8: The East Tower was inspected using an aerial robot: (a) view of the joints between the tower column, slab, and web; (b) magnified, lightened, and clarified detail showing opened cold joint between tower column and slab; and (c) magnified, lightened, and clarified detail showing large crack in the web Vertical component of force imposed by stays anchored to abutment = 3600 kN

B

Line of action of stays anchored to abutment 4200 kN

Mast weight = 12,000 kN Upper column weight = 2(8000 kN)

98.5 m

6.5 kPa(13 m) = 84.5 kN/m

40° 7

6

5

4

3

2 1

1

2

3

4

5 6

7

8

9 10 11 12 13 14

1110 kN 129.5 m

66 m

Lower column and web weight Foundation reactions

Fig. 9: A free-body diagram of the West Tower and deck at the time of the collapse (Note: 1 m = 3.3 ft; 1 kN = 225 lbf; 1 kPa = 21 psf)

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Linear idealization of the tower and web

We also made a linear idealization of the tower, including a membrane to represent the web joining the columns of the lower half of the tower. This model resulted in a tensile force in the critical tower tie of about 1.5 million lbf (6670 kN). Again, this force exceeds the capacity of the tower slab. The number of strands required to resist this tension (working at strength) is 26 (many more than the 12 strands provided). Including factors of safety and accounting for loads not acting at the time of collapse, one would expect this number to nearly triple (to ~78 strands). Interestingly, practically zero tension would be expected in the transverse direction of the tower slab (longitudinal direction of the deck). Nevertheless, the slab drawings show 14 tendons, each containing eight strands, for a total of 112 strands in that direction.

Finite element model of the tower and web

What is more difficult to explain is why the West Tower supported loads for as long as it did and why the East Tower remained standing after the collapse of the West Tower. We believe that failure did not occur earlier during construction because the structure was able to accommodate some force redistribution through cracking and yielding of reinforcement in the web of the tower. To explore the issue further, we developed an approximate model of the tower using SAP2000,2 based on the following assumptions: The tower was idealized as frame elements, with moments transferred at column joints but with no moment or axial load transferred at the connection between the columns and the tower slab. In other words, the cold joints between the slab and columns were represented by axial and moment releases, uncoupling the corresponding degrees of freedom between the slab and columns; The tower slab and columns were modeled using linear beam elements, with an assumed elastic modulus of 25,000 MPa (3.6 million psi); The tower web between the lower columns was modeled using nonlinear layered shell elements, with reinforcing bars having a yield stress of 480 MPa and concrete having a tensile strength of 1.7 MPa (250 psi). The tower slab acted as a stiffener at the web top; and The force applied by the 0.6 in. diameter unbonded strands in the tower slab was modeled as an external clamping force. Three analyses were made using different clamping forces. The lowest force was 3100 kN, which is approximately the

••

•• •• ••

(a)

(b)

tensile strength of the 12 unbonded strands in the tower slab. The other clamping forces were 3560 and 4450 kN. Analyses were made in U.S. customary units (clamping forces of 700,000; 800,000; and 1 million lbf). At the highest clamping force, no yielding was calculated in the horizontal web reinforcement. For the lower clamping force values of 3100 and 3560 kN, yielding and concentrated strains were calculated at the column-web joint. Figure 10 shows horizontal reinforcement stress for the three analyses. Note that stresses in the slab were also observed to be below the expected tensile strength of concrete in direct tension (which can be as low as 1.4 to 2.1 MPa [200 to 300 psi]). Strain concentration, leading to brittle response, could be expected at the web-column connection because the cracking stress of the concrete in direct tension was about the same as the gross stress over the concrete section at yield of the horizontal reinforcement in the web. With a horizontal web reinforcement ratio of 0.26%, the gross stress in the concrete can be calculated as 0.26% × 480 MPa = 1.25 MPa (180 psi). This condition will result in limited cracking, as strain will be concentrated at initial cracks. Further, the unreinforced cold joint at the slab ends would have acted as a crack initiator. Heavier web reinforcement would have produced a more ductile structure and possibly provided more time to react to the failure. Cracking observed in the remaining (East) tower (shown in Fig. 8) supports the hypothesis that strains concentrated at the cold joint at the slab level and near the web-column connection. Why did the remaining tower not collapse? Here are a few plausible reasons: Because concrete had not yet been placed in the two most

••

(c)

Fig. 10: Plots of stress in horizontal reinforcement in the web, determined using a finite element model of the tower (the full tower was modeled, but columns are truncated in the illustration). Stress in horizontal reinforcement at the column-web joint varied significantly with clamping force: (a) clamping force P = 700,000 lbf (3100 kN); (b) P = 800,000 lbf (3560 kN); and (c) P = 1 million lbf (4450 kN) www.concreteinternational.com | Ci | JUNE 2019

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distant segments in the cantilever of the East Tower, the East Tower had lower loads. The 200 mm thick concrete deck resulted in a uniform load of 4.7 kPa applied over the 13 m wide deck area. Thus, relative to the West Tower, the total stay force in the cantilever side could have been reduced by 61 kN/m × 2 × 9.25 m = 1100 kN. Further, the horizontal force in the stays anchored to the abutment could have been reduced by 1340 kN to balance the cantilever, resulting in a plausible reduction in the stay force in the tower of 1100 + 1340 × tan(40) = 2220 kN (a 10% reduction); Post-tensioning in the tower slab of the East Tower might have been more effective than in the collapsed West Tower; The East Tower might have had more web reinforcement; or Initial shrinkage cracks might have distributed strains better and reduced brittleness in the standing tower. The answer may involve one or several of these plausible reasons. In any case, it is clear that the East Tower was near its limit. (Following a detailed study by a consulting firm, the remaining portion of the Chirajara Bridge was brought down

•• •• ••

by a controlled detonation in July 2018.) The concentration of strains described previously implies that the wall web reinforcement was likely to fracture early in the collapse process. That failure would not necessarily lead to collapse if the remaining structure had sufficient capacity. A limit analysis of a failure mechanism that did not include web yielding provided an estimated resistance smaller than the estimated demand. For safe operation, the demand should be much less than the resistance. Given the brittleness of the web, it is unlikely that both web and columns (working as parts of a plastic mechanism) reached their capacities at the same time.

Conclusions

In our opinion, the West Tower of the Chirajara Bridge collapsed because post-tensioned reinforcement placed in a slab meant to act as a tie between columns was insufficient by a large margin. In contrast, in the perpendicular direction (that is, in the direction of the deck), where no large stresses would have been expected, this slab tie (or tower slab) was provided with nine times more reinforcement. Had the provided

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reinforcement been rotated 90 degrees, the collapse would not have occurred. The insufficient lateral restraint offered by the reinforcement in the tower slab—a critical component that acted as a tie between columns—forced the reinforced concrete web between the columns to work in tension. Two factors played a critical role in the response of the web to this tension: The web had minimum tensile reinforcement (with a horizontal web reinforcement ratio of 0.26%). Its strength was comparable to the expected cracking stress of the concrete in direct tension. When cracking stress and reinforcement strength are similar, deformations concentrate at a single or few cracks and tend to result in brittle failure; and The cold joint formed at the connection between the tower slab and the columns created a discontinuity, forcing further concentration of strain and increasing the brittleness of the system. Critical elements required to work in tension should be reinforced with amounts of reinforcement exceeding the code minima to avoid brittleness. The brittleness of the web prevented better redistribution of forces, which could have delayed the collapse. This tragic example emphasizes once again the importance of peer review of all projects—especially projects with large and complex structures.

••

••

Acknowledgments Our deepest condolences are expressed to the families of the workers who lost their lives in this tragedy. The only purpose of our work is to avoid similar tragedies in the future. Xenital S.A.S., under the direction of Sebastian Uribe, assisted the team in the inspection of the bridge using a drone-mounted camera. The Agencia Nacional de Infraestructura (ANI), Fiscalia General, and Ministerio de Transporte, all government agencies in Colombia, are thanked for their assistance in our investigation. Special gratitude is expressed to Ron Burg and Matthew Senecal of ACI, Luis Fernando Mejia of ANI, Luís E. Garcia, and Omar D. Cardona. Their support and assistance were indispensable.

Santiago Pujol, FACI, is a Professor of civil engineering at the Lyles School of Civil Engineering at Purdue University, West Lafayette, IN. His research interests include seismic vulnerability of structures, response of structures to impulsive loads, structural health monitoring, repair, and strengthening of structures. He is a member of ACI Committees 133, Disaster Reconnaissance, and 314, Simplified Design of Concrete Buildings; ACI Subcommittee 318-R, High Strength Reinforcement; and Joint ACI-ASCE Committees 441, Reinforced Concrete Columns, and 445, Shear and Torsion.

Michael E. Kreger, FACI, is the Garry Neil Drummond Endowed Chair in Civil Engineering and Director of the Large-Scale Structures Laboratory in the Department of Civil, Construction and Environmental Engineering at the University of Alabama, Tuscaloosa, AL. He is a member of the ACI Board of Direction and ACI Committees 133, Disaster Reconnaissance; 318, Structural Concrete Building Code; and 374, Performance-Based Seismic Design of Concrete Buildings; ACI Subcommittees 318-C, Safety, Serviceability, and Analysis; and 318-J, Joints and Connections; and Joint ACI-ASCE Committees 352, Joints and Connections in Monolithic Concrete Structures, and 441, Reinforced Concrete Columns.

Jonathan D. Monical is pursuing his PhD at the Lyles School of Civil Engineering at Purdue University, where he received his BSCE in 2016 and his MSCE in structural engineering in 2017. He is a recipient of the school’s 2017 John E. Goldberg Fellowship.

Disclaimer This article contains the opinions and observations solely of the writers. It does not reflect the views of any organization with which they are associated. The writers shall not assume any responsibility related to use of their opinions by others.

References 1. ANI website, Agencia Nacional de Infraestructura (National Infrastructure Agency), Colombia, www.ani.gov.co/proyecto/carretero/ bogota-villavicencio-21255. (last accessed Feb. 2019) 2. SAP2000, “Integrated Software for Structural Analysis and Design,” Computers & Structures, Inc. (CSI), 2018.

Arturo E. Schultz is a Professor of civil engineering and Director of Hybrid Simulation at the Department of Civil, Environmental, and Geo- Engineering at the University of Minnesota, Twin Cities, Minneapolis, MN. His research interests include the seismic design and performance of concrete and masonry structures. He is a member of Joint ACIASCE Committees 335, Composite and Hybrid Structures, and 441, Reinforced Concrete Columns.

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